Insitu Testing

Introduction

In terms of foundation engineering, insitu (or in place) testing refers to those methods that evaluate the performance of a geotechnical structure or the properties of the soils or rock used to support that structure. This testing can range from a soil boring at a surveyed location to monitoring the response of a fully loaded bridge pier, dam, or other foundation element. The reliability of a given structure to function as designed is directly dependent on the quality of the information obtained from such testing. Therein, it is imperative that the design engineer be familiar with the types of tests and the procedures for proper execution as well as the associated advantages and disadvantages.

Methods of insitu evaluation can be invasive or noninvasive, destructive or nondestructive, and may or may not recover a specimen for visual confirmation or laboratory testing. Invasive tests (e.g. soil borings or penetration tests) tend to be more time consuming, costly, and precise; whereas, noninvasive tests (e.g. GPR or siesmic refraction) provide a large amount of information in a short period of time that is typically less quantifiable. However, when used collectively, the two methods can complement each other by: (1) defining areas of concern from noninvasive techniques, and (2) determining the foundation design parameter from invasive techniques. This is particularly advantageous on large sites where extreme variations in soil strata may exist.

Site Exploration Plan

The subsurface investigation program for a given site should account for the type, size, and importance of the proposed structure. These parameters help focus the design of the site exploration program by determining the quantity and depth of soil soundings (or borings) required. Planning for a site investigation should also include the location of underground utilities (i.e. phone, power, gas, etc.). As such, a local "call before you dig" service should be notified several days prior to the anticipated investigation. These services are usually subsidized by the various local utilities and have no associated cost.

The quantity of borings is largely dependent on the overall acreage of the project, the number of foundations, or the intended use of the site. For foundations, the depth of borings depends on the zone of soil influenced by the foundation geometry and the given loading. For instance, a proposed roadway alignment typically requires a hand-auger investigation every 100 feet along the centerline to a depth of 5 feet to define uniformity of the subgrade material as well as spatial variability. Therein, the importance of the structure, in the form of causal effects should a failure occur, is somewhat minimal. Further, if undesirable soils conditions were identified, a followup investigation could be requested. In contrast, preliminary borings along the alignment of a proposed bridge foundation can be more frequent and are much deeper depending on the depth to a suitable bearing strata. At a minimum, one boring should be performed at each pier location to a depth 3 - 5 foundation diameters below the anticipated foundation. Likewise, buildings with large column loads often require a boring at each column location unless the soil shows extremely consistent behavior. For extremely important structures the designer and/or client not only require more scrutiny from the subsurface investigation, but also require an amplification factor (or importance factor) be applied to the load to assure a low probability of failure (AASHTO, 1997).

In virtually all cases, the additional cost of more thorough subsurface investigation can be reconciled with a more cost effective foundation design. Uncertainty in subsurface conditions in most instances promotes needless over-design. Depending on the type of design to be considered, the designer must recognize the effect of site variability as well as the type of testing that can be conducted to increase confidence and reduce the probability of failure.

Geophysical Testing Methods

Ground Penetrating Radar

Ground Penetrating Radar (GPR) is a geophysical exploration tool used to provide a graphical cross-
section of subsurface conditions. This cross-sectional view is created from the reflections of repeated short-duration electromagnetic (EM) waves which are generated by an antenna in contact with the ground surface as the antenna traverses across the ground surface. The reflections occur at the interfaces between materials with differing electrical properties. The electrical property from which variations cause these reflections is the dielectric permittivity, which is directly related to the electrical conductivity of the material. GPR is commonly used to identify underground utilities, underground storage tanks, buried debris, or geological features. The information from GPR can be used to make recommendations for more invasive techniques such as borings. Figure 1 shows a ground launch GPR system being pushed along a predetermined transect line.

The greater the electrical contrast between the surrounding earth materials and the target of interest, the greater the amplitude of the reflected return signal. Unless the buried object/target of interest is highly conductive, only part of the signal energy is reflected back to the antenna located on the ground surface with the remaining portion of the signal continuing to propagate downward to be reflected by deeper features. If there is little or no electrical contrast between the target of interest and the surrounding earth materials, it would be very difficult if not impossible to identify the object using GPR.

The GPR unit consists of a set of integrated electronic components which transmits high frequency (100 to 1500 megahertz [MHz]) electromagnetic waves into the ground and records the energy reflected back to the ground surface. The GPR system is comprised of an antenna, which serves as both a transmitter and receiver, and a profiling recorder that both processes the data and provides a graphical display of the data.

The depth of penetration of the GPR is very site specific and is controlled by two primary factors: subsurface soil conditions and antenna frequency. The GPR signal is attenuated (absorbed) as it passes through earth materials. As the energy of the GPR signal is diminished due to attenuation, the energy of the reflected waves is reduced, eventually to a level where the reflections can no longer be detected. In general, the more conductive the earth materials, the greater the GPR signal attenuation. In Florida, typical soil conditions which severely limit the GPR signal penetration are near-surface clays, organic materials, and the presence of sea water in the soil pore water space.

A GPR survey is conducted along survey lines (transects) which are measured paths along which the GPR antenna is moved. Known reference points (i.e., building corners, driveways, etc.) are placed on a master map, which includes traces of the GPR transects overlying the survey geometry. This survey map allows for correlation between the GPR data and the position of the GPR antenna on the ground.

For geological characterization surveys, the GPR survey is conducted along a set of perpendicularly oriented transects. The survey is conducted in two directions because subsurface features are often asymmetric for residential surveys. Spacing between the survey lines is initially set at 10 feet. More closely spaced grids may be used to further characterize a recorded anomaly.

The features observed on the GPR data that are most commonly associated with potential sinkhole activity are:

A down-warping of GPR reflector sets, that are associated with suspected lithological contacts, towards a common center. Such features typically have a bowl or funnel shaped configuration and are often associated with deflection of overlying sediment horizons caused by the migration of sediments into voids in the underlying limestone (Figure 2). In addition, buried depressions caused by differential subsidence over buried organic deposits and debris may also cause these observed features.

A localized significant increase in the depth of penetration and/or amplitude of the GPR signal response. The increase in GPR signal penetration depth or amplitude is often associated with a localized increase in sand content at depth.

An apparent discontinuity in GPR reflector sets that are associated with suspected lithological contacts. The apparent discontinuities and/or disruption of the GPR reflector sets may be associated with the downward migration of sediments.

The greater the severity of the above features or a combination of these features, the greater the likelihood that the identified feature is related to past or present sinkhole activity.

Depth estimates to the top of the lithological contacts or targets of interest are derived by dividing the time of travel of the GPR signal from the ground surface to the top of the feature by the velocity of the GPR signal. The velocity of the GPR signal is usually obtained for a given geographic area and earth material from published sources. In general, the accuracy of the GPR-derived depth estimates ranges from ±25 percent of the total depth.

Although the GPR is very useful in locating significant lithologic soil changes, strata thickness, and inferred subsurface anomalies, the GPR cannot identify the nature of earth materials nor their condition (i.e., loose vs. dense sand, soft vs. stiff clay). The GPR data is best used in conjunction with other geotechnical and physical tests to constrain the interpretation of the virtual cross-section profiles.

Resistivity Tests

Electrical Resistivity Imaging (ERI) is a geophysical method that maps the differences in the electrical properties of geologic materials. These changes in electrical properties can result from variations in lithology, water content, pore-water chemistry, and the presence of buried debris. The method involves transmitting an electric current into the ground between two electrodes and measuring the voltage between two other electrodes. The direct measurement is an apparent resistivity of the area beneath the electrodes that includes deeper layers as the electrode spacing is increased. The spacing of electrodes can be increased about a central point, resulting in a vertical electric sounding (VES) that is modeled to create a 1-D geoelectric cross-section. Recent advances in technology allow for rapid collection of adjacent multiple soundings along a transect that are modeled to create a 2-D geoelectric pseudo cross-section. The cross-section is useful for mapping both the vertical and horizontal variations in the subsurface (see Figure 3).

Although the result from this method are not absolute, the resistivity trends obtained are useful for mapping stratigraphy such as aquatards, bedrock, faults and or fractures. It can delineate anomalous formations or voids in karstic material, the presence of salt water intrusion in coastal regions, and detect leaks in dams as well as other applications. It is most successful in large cleared areas without severe changes in topography; not recommended for small congested or urban sites. Buried utilities or other highly conductive anomalies can adversely affect the results.

This method is fast, non-invasive, and relatively inexpensive when compared to drilling. When compared to electromagnetic methods, is it less susceptible to interference from overhead power lines. It is easily calibrated to existing boreholes to allow for correlations between measured resistivity and estimated soil properties. As with other geophysical test methods, it is best suited for environmental or water resources disciplines that require stratigraphy or soil property mapping of large land parcels.

Seismic Refraction

The seismic refraction technique measures the seismic velocity of subsurface materials and models the depth to interfaces with a velocity increase. Soil conditions and geologic structure are inferred from the results, since changes in material type, or soil conditions, are often associated with changes in seismic velocity. Seismic energy, which is introduced into the subsurface using a drop weight or explosive source, propagates through the earth as a wave front that is refracted by the material through which it passes. As illustrated in Figure 4, the wave front intersects a high-velocity interface, creating a "head wave" that travels in the higher velocity material nearly parallel to the interface. The energy in this head wave leaves the interface and passes back through the low velocity material to the surface. Geophones placed at selected intervals along the ground surface detect the ground motion and send an electrical signal, via a cable, to a recording seismograph. The objective is to determine the arrival times of these refracted waves in order to calculate the velocity of the material and model the depth to various interfaces.

This test is used to determine stratigraphy such as the depth to bedrock. It is best suited for stratigraphy that increases in density (or seismic velocity) with depth. In such cases it can estimate the depth of borrow materials, aid in mapping faults and fractured rock zones, locate the water table, estimate material elastic properties such as shear modulus and Poisson's ratio. The depth of exploration is limited by the energy source and the maximum length of geophone spacing. The test is less expensive most soil exploration methods as well as other comparable seismic reflection methods. The vertical resolution is usually better than electrical, magnetic or gravity methods of site investigation.

Physical Sampling and Penetration Tests

Standard Penetration Test

The standard penetration test (SPT) is undoubtedly the most common method of soil exploration for foundation design. It is an invasive test that not only provides information from which soil strength can be estimated, but also it provides a physical sample that can be visually inspected or used for laboratory classification. Although the test method has undergone several iterations with respect to upgrading equipment, it is sensitive to operator and equipment variability. Regardless, the general concept of penetration resistance and the hands on soil sample recovery make it the choice of many designers.

The SPT is described by the American Society for Testing and Materials (ASTM) as test number D-1586, entitled "Standard Method for Penetration Test and Split-Barrel Sampling of Soils." This standard defines the appropriate manner in which the test should be conducted which involves drilling techniques, penetration and sampling methods, proper equipment, and the reporting of results. In general, a 2" O.D. split spoon sampler is driven into the ground with a 140 lb drop hammer dropped 30 inches repeatedly until a penetration of 18 inches is achieved . The number of blows of the hammer is recorded for each of three 6" intervals (totaling 18 inches). The number of blows required to advance the sampler the last 12 inches (2nd and 3rd interval) is defined as the SPT N-value. Upon extraction of the sampler, the soil retrieved is visually inspected, documented, and placed in jars for more elaborate testing (if so determined by the engineer). At best, continuous sampling produces a single SPT N-value every 1.5 ft. At minimum, a sample should be taken every 5 feet of depth.

Between each penetration test, a boring should be advanced to permit the next sample without interference from side shear resistance along the length of the drill rod. Several boring techniques are acceptable: one-hole rotary drilling, continuous flight hollow stem augering, wash boring, or continuous flight solid stem augering. However, under no circumstances should the soil beneath the advanced borehole be disturbed by jetting or suction action caused by improper drilling techniques. For instance, extracting a continuous flight auger from submerged soils will reduce the insitu stresses and produce lower N-values.

The hammer used to drive the sampler can either be manual or automatic. Numerous configurations of both hammer types have been manufactured. The safety type is the most common manual hammer as it is equally suited to both drive and extract the split spoon. This type of hammer is lifted by the friction developed between a rope and a spinning cathead power take-off. The number of wraps around the cathead as well as the diameter of cathead are specified as well as the condition of the rope and cathead surface (Figure 5). Due to the incomplete release of the drop weight from the cathead, the total potential energy of the drop is not available to advance the sampler. A recent study showed that manual hammers transfer anywhere between 39 and 93% of the energy (average 66%), while automatic hammers ranged between 52 and 98% (average 79%). Although the reproducibility of a automatic hammer is better than manual hammers, the variation in energy efficiency cited is dependent on the upward velocity of the hammer as controlled by the revolutions per minute of the drive chain motor (Figure 6). To this end, a given machine should be calibrated to produce an exact 30 inch drop height and the rpms required to produce that drop recorded and maintained.

Aside from the visual and physical classification that can be obtained from a standard penetration test, correlations have been established that provide estimates of insitu soil properties based on the soil type and blow count. These correlations can be based on the corrected or uncorrected SPT blow count N' or N, respectively.

Corrected blow counts provide a method of accounting for the insitu state of stress surrounding a soil sample while it was being tested. For instance, sands with identical structure with appear stronger (higher blow counts) at greater depths than when at shallower depths. As such, soil properties such as unit weight may be better estimated if overburden effects are removed or normalized. However, soil properties such as shear strength or available end bearing are enhanced by greater insitu stresses and are generally correlated to uncorrected blow counts. The following expression is used to correct SPT N-values by normalizing it to a 1 tsf overburden insitu state.

Tables SPT-1 and SPT-2 provide estimated values for corrected and uncorrected blow counts, respectively.

Table SPT-1 (FHWA, 1993)

Corrected

SPT-N'	Description	moist
(pcf)	(kN/m3)	(deg)
SANDS
0	Very Loose	70 - 100	11.0 - 15.7	25-30
4	Loose	90 - 115	14.1 - 18.1	27-32
10	Medium	110 - 130	17.3 - 20.4	30-35
30	Dense	120 - 140	18.8 - 22.0	35-40
50	Very Dense	130 - 150	20.4 - 23.6	38-43
CLAY	qu, ksf
0	Very Soft	100 - 120	15.7 - 18.8	0
2	Soft	0.5
4	Medium	110 - 130	17.3 - 20.4	1.0
8	Stiff	2.0
16	Very Stiff	120 - 140	18.8 - 22.0	4.0
32	Hard	8.0

Table SPT-2 (Kulhawy and Mayne, 1990)

SPT-N	sat	sub
(pcf)	(kN/m3)	(pcf)	(kN/m3)	(deg)
SANDS
0-2	100	15.7	37.6	5.9	26
3-4	100	15.7	37.6	5.9	28
4-10	105	16.5	42.6	6.7	29
10-20	110	17.3	47.6	7.5	30
20-30	115	18.1	52.6	8.3	32
30-40	120	18.9	57.6	9.1	33
>40	125	19.6	62.6	9.8	34
CLAY
0-2	105	16.5	42.6	6.7	0
2-4	110	17.3	47.6	7.5	0
4-8	115	18.1	52.6	8.3	0
8-15	120	18.9	57.6	9.1	0
15-30	125	19.6	62.6	9.8	0
>30	125	19.6	62.6	9.8	0
Clay Shear Strength C = N / Ti in psf, where Ti is the soil type factor

Ti = 8 for most clay

Ti = 10 for low plasticity or SC

Ti = 12 for peat
LIMESTONE
SPT - N	Shear Strength

sat = 135 pcf (21.2 kN/m3)

sub = 72.6 pcf (11.4 kN/m3)

= 0 deg

Ka = 1.0

Kp = 1.0
psf	kN/m2
10-20	4000	190
20-50	8000	380
50-100	15000	720

Insitu Rock Testing

The design of rock-socketed drilled shafts is highly dependent on the integrity of the rock core samples obtained from field investigation. When sufficient samples are recovered, laboratory tests can be conducted to determine the splitting tensile strength, qt, (ASTM D 3967) and the unconfined compressive strength, qu,(ASTM D 2938). The shear strength of the shaft/limestone interface, fsu , is then expressed as a function of qt and qu (McVay et al., 1992). This value is typically adjusted by rock quality indicators such as the Rock Quality Designation, RQD, (ASTM D 6032) and/or the percent recovery, REC. For example, the State of Florida outlines a method using the percent recovery to offset the highly variable strength properties of the Florida limestone formation. Therein, a design value, (fsu)DESIGN, is expressed as REC * fsu (Lai, 1999). These methods work well in consistent, competent rock but are subject to coring techniques, available equipment, and driller experience. Sampling problems are compounded in low quality rock formations as evidenced by the occurrence of zero RQD and low REC values.

Timed Drilling. To counter poor quality samples (or no sample at all) some designers with extensive local experience use timed drilling techniques to estimate rock quality and shaft design values in addition to, or in lieu of, the previous methods. With this technique, the driller must record the time to advance a wash boring through a bearing strata while maintaining a constant "crowd" pressure, fluid flow, and rotational bit speed. Advance times would typically need to be greater than 2 - 3 minutes per foot to be useful. Lower advance times are common in weaker soils which are more effectively tested by Standard Penetration Testing. Like SPT and CPT, the equipment should maintain reasonably consistent physical dimensions (i.e. the bit should stay in good working condition). Although this method is very simple, it is highly empirical and largely dependent on the uniformity of the drilling techniques. Additionally, the designer must have developed a large enough database (with load test calibration) to design with confidence. Such databases exist, but are proprietary and not common knowledge.

Coring Methods. When designing from rock core samples it is important to consider the factors affecting sample retrieval and hence their quality. The recovered samples can range in diameter from 0.845" to 6" where larger samples are preferred in soft limestone. The State of Florida requires a minimum core diameter of 2.4" but recommends 4". The drill core samples can be obtained from three different types of core barrels: single tube, double tube, and triple tube. The simplest is the single tube in which the drill core and flushing fluid occupy the same space and consequently can lead to erosion of low strength or fragmented rock samples. As a result, this type of core barrel is not permitted for use with Florida limestone (FDOT, 1999).

Double tube core barrels differ from single tube by essentially isolating the drill core from the flushing fluid (see Figure 7). Simple versions of this type of core barrel use a rotating inner tube which requires a small fraction ( approx. 10%) of the drilling fluid to circulate around the drill core to prevent binding and direct contact of the sample with the tube. Most double tube systems now use a fixed inner tube which requires no flush fluid around the drill core and thus causes less disturbance to the sample. During extraction of the entire barrel assembly, a core trap-ring at the leading edge of the inner barrel snares the drill core preventing its loss (see Figure 8). Recovering the sample from the inner tube without disturbing it is difficult in soft, fragmented, or inter-layered rock deposits. Both fixed and rotating inner core barrels are permitted by FDOT but significant variations in recovery values should be expected.

The triple tube core barrel, in concept, is essentially the same as the double tube (with the fixed
inner tube). It differs in the way the specimen is recovered in that the inner tube is fitted with yet a third sleeve or split tubes in which the drill core is housed. The entire sleeve or split tube is extruded from the inner barrel using a plunger and pressure fitting that pushes directly on the split tubes. The extrusion process is similar to that of Shelby tube samples except the sample is not stressed. In this manner, the sample is not compressed or shaken loose. Figures 9-12 show the components of the triple tube core barrel and sample extruder.

Further variables affecting core drilling results include: the type of drill bit, the flow rate of the flushing fluid, the end gap between the inner and outer barrels, the crowd pressure, and the advance rate through softer inter-layered deposits. With so many variables controlling sample recovery, methods of investigating the remaining borehole for the insitu limestone characteristics could have significant merit.

Insitu Rock Strength Tests. Direct measurements of the insitu bond/shear strength of the drilled shaft-to-rock interface can be obtained through small-scale anchor pull-out tests or full-scale load tests. Anchor pull-out tests are purported to have produced reasonable correlations with full-scale results (Bloomquist et al., 1991). The test method involves simply grouting a high strength post tensioning rod into a borehole, and measuring the load required to pull the grout plug free (Note: load is directly applied to the base of the plug to produce compression and the associated Poisson expansion in the specimen). Attention must be given to the surface area formed by the volume of grout actually placed. This test is an attractive option in that it is relatively inexpensive, requires minimal equipment mobilization, and can be conducted at numerous locations throughout a site. However, it has received little attention as a whole and remains comparatively unused.

Design-phase, full-scale, insitu testing of the shaft/limestone interface is by far the surest method to determine the design parameters of a drilled shaft. This can be accomplished by several means: top down static loading, bi-directional static loading, statnamic loading, or drop-hammer dynamic loading (discussed later). However, due to the associated costs, only a small fraction of rock sockets are tested in this fashion, and rarely at the design phase. Additionally, a single test may not adequately account for the spatial variability of rock formations without correlation to standard site investigation methods. As such, a host of insitu borehole devices have been developed to aid in estimating soil and rock strength parameters.

Insitu borehole modulus devices are classified into two categories based on their loading apparatus: (1) rotationally symmetric borehole loading devices, and (2) diametrically arranged lateral loading plates (Wittke, 1990). Figure 13 shows the loading scheme of the two conceptual mechanisms.

Type (1) probes apply load via a rubber diaphragm which is pressurized by either gas or liquid. In general, measurements of displacement are made directly when using gas pressure, and indirectly through change in volume when using fluid pressure. Table 1 lists Type (1) devices that have been developed by various manufacturers.

Type (2) probes use two semi-cylindrical thrust plates diametrically aligned to apply loads to the arc of the borehole. Measurements of displacement are obtained directly at a minimum of two locations along the longitudinal axis of the plates. Whereas Type (1) devices produce uniformly distributed radial stresses in the borehole, the stress distribution in Type (2) devices is dependent on the relative stiffness of the rock and the plate. Table 2 lists Type (2) devices that have been developed by various manufacturers.

Table 1 List of Type (1) Borehole Devices (after Wittke, 1990)
Name	Method of Measuring	Number of Measuring Devices	Max. Applied Pressure Pmax [MN/m2]	Borehole Diameter d[mm]	Test length l [mm]	l/d
Menard pressiometer	indirect ( v)	-	10	34-140	502-910	≥65
CSM cell	indirect ( v)	-	70	38	165	4.3
Janod Mermin probe	direct ( d)	3	15	168	770	4.6
Sounding dilatometer	direct ( d)	2	4/7.5	200/300	1000/1200	5/4
Comes probe	direct ( d)	3	15	160	1600	10
LNEC dilatometer	direct ( d)	4	15	76	540	7.1
Tube deformeter	direct ( d)	4	4	297	1300	4.4
Prigozhin pressiometer	direct ( d)	2	20	46	680	14.8
Atlas dilatometer	direct ( d)	8	10	144	890	6.2
BGR dilatometer	direct ( d)	4	40	86	1000	11.6
Dilatometer 95	direct ( d)	3	12	100	1000	10
Dilatometer 112	direct ( d)	3	12	116	1000	8.6
Elastometer 100	direct ( d)	2	10	62	520	8.4
Elastometer 200	direct ( d)	3	20	62	520	8.4

Table 2 List of Type (2) Borehole Devices (after Wittke, 1990)

Name	Thrust plates' angle of opening	Max. applied pressure Pmax [MN/m2]	Borehole diameter d [mm]	Test length l [mm]	l/d
Geoextensometer	2 = 143°	34	76	306	4.0
Goodman jack	2 = 90°	64	74-80	204	2.6-2.8
CSIRO pressiometer	2 = 120°	35	76	280	3.7

Another mechanism (not originally intended for rock) that has interesting features in respect to weak rock is the Iowa borehole shear device. The test scheme for this device is a combination of both the anchor pull-out test and the borehole modulus test. The device is expanded into the walls of the borehole and is then pulled to determine the shear strength of the soil. Typically, several lateral pressures are investigated (Demartinecourt, 1985).

Cone Penetration Test

The cone penetration test (CPT) is an invasive soil test that defines soil strata type, soil properties, and strength parameters. It is highly repeatable, insensitive to operators, and best suited for uncemented soils, sands, or clay. Although this test retrieves no sample for laboratory testing or visual inspection, it has the capability of producing enormous amounts of physical information based on correlations with side-by-side tests with other test methods such as SPT. Further, as the test provides direct measurements of ultimate end bearing and side shear, it is directly applicable for design of foundations of all kinds.

The CPT is described in ASTM test number D-3441, entitled "Standard Test Method for Deep, Quasi-Static, Cone and Friction-Cone Penetration Tests of Soil." This is to include cone penetration type tests that use mechanical or electronic load detection, tip or tip and friction stress delineation, and those tests where the penetration into the soil is slow and steady in a vertically aligned orientation. Those tests conducted with mechanical load detection are typically denoted as Dutch Cone Tests and those using electronic detection are simply called Cone Penetration Tests. The term "quasi-static" refers to a steady rate of penetration where the acceleration is zero, but the velocity of penetration is constant (1 - 2 cm/sec ± 25%).

The test apparatus consists of a 60 degree conical tip of known cross-sectional area that is thrust into the soil at a near constant rate. Behind the cone tip, a friction sleeve of known surface area is also included that is used to detect the side shear or adhesion between the steel sleeve and the surrounding soil. The force required to advance the tip through the soil is divided by the cross-sectional area to determine the tip stress, qc. Similarly, the force required to advance the friction sleeve is divided by the sleeve surface area to produce the local friction value, fs. The tip area and sleeve area vary from device to device but the most common areas are 10 cm2 and 150 cm2 , respectively.

The tip area (diameter) can influence the magnitude of the resulting qc value similar to the effects of foundation diameter on capacity. This is due to the increased zone of influence beneath the tip as the cone diameter increases for various devices. Therefore, in relative uniform soils, the tip diameter has little effect. In layered or more heterogeneous strata, a smaller tip diameter will better register the minute changes in soil type and strength. Larger diameter cones physically average the effects of thin layers. Figure 14 shows two different-sized cone tip and sleeve assemblies.

Another feature that CPT equipment usually incorporate is the capability of monitoring the pore water pressure while advancing the cone/sleeve assembly. There can be a significant amount of excess pore pressure developed by forcing the volume of the cone into a somewhat fixed volume of a poorly draining material. In contrast, in dense fine-grained soils, the cavity expansion can cause a decrease pore pressure. The assemblies shown in Figure 14have a pressure transducer within the tip body that registers the pore pressure directly behind the tip (between the tip and sleeve). This information can be used to convert the total stress registered by the tip to effective stress similar to a consolidated undrained triaxial compression test. When pore pressure measurements are taken the test is denoted as CPTU. Smaller cones tend to induce less cavity expansion and therefore less effects on total stress.

The thrust required to advance the cone assembly is dependent on the strength of the soil as well as the size of the tip. Given the disparity between the cone sizes in Figure 14, it is not surprising that the size of the equipment required to advance these cones is also disparate. Figure 15 shows the associated truck mounted CPT rigs that use these devices. The larger diameter cone requires a 20 ton thrust mechanism and can reach depths of 100 meters; the mini-cone requires about 1/5 the thrust and can be mounted on the front of a standard truck or utility vehicle. Mini-CPTs are limited to a practical maximum depth 20 meters.

The thrust mechanisms also vary between the two systems in Figure 15. The standard CPT system uses 1 meter segmental rods to advance the cone tip/sleeve assembly. The thrusting ram is designed to grip, thrust downward 1 meter, release and stroke upward 1 meter, re-grip and repeat. At a penetration rate of 2 cm / sec, the process progresses at an average advance rate of 1 cm / sec to account for the re-gripping. An average sounding to 30 meters should take on the order of 1.5 - 2 hours (in and out). A SPT to a similar depth could take multiple days. The mini-cone uses a more time efficient method of advancing the cone. It uses a contiguous length of mild stainless steel tubing (10 - 20 meters long) that is continuously feed at a constant rate without having to re-stroke and re-
grip. Therein, the tubing is straightened from a large diameter coil as it is continuously gripped by side-by-side opposing chains specially designed to mate to the coil diameter. Figure 16 shows both thrust mechanisms.

The data collection during cone penetration testing is typically taken at 5 cm intervals but can be taken as frequent as permitted by the data acquisition system. This gives a virtually continuous sounding of tip and sleeve stresses. As both of these devices use strain gage based load cells, the instrumentation leads are routed through the center of the thrusting rod / tube. The data are processed to produce the soil type as well as other parameters in real time. The basis for the data regression is based on correlations developed by Robertson and Campanella, (1983). Although many correlations exist, the most significant uses a calculated parameter that defines the ratio of measured side to the measured tip stress. This ratio is defined as the friction ratio, Rf. To aid in classifying various soils, twelve soil types were defined that could be readily identified given the cone bearing stress, qc, and the friction ration, Rf. Figure 17 shows the classification chart used to identify soils from CPT data. Further, correlations from CPT to SPT test data were developed to elevate the comfort of designers more familiar with SPT data. Therein, the qc/N ratio was defined for each of the twelve soil types (also shown in Figure CPT 7).

Quality Assurance Test Methods

The construction of foundation is plagued with unknowns associated with the integrity of the as-built structure. This is particularly problematic with deep foundations that are installed without visual certainty of the actual conditions or configuration. This section will discuss several methods used to raise the confidence of the design with regards to concrete quality or capacity verification.

Shaft Inspection Device (SID)

The inspection device is a visual inspection system for evaluating bottom cleanliness of drilled shaft excavations. A special video camera contained in a weighted, trapped-air bell housing is lowered into the shaft excavation prior to concreting to record the condition of the bottom. This is particularly helpful in slurry excavations where quality assurance is difficult to maintain. The bell housing is outfitted with gages in clear sight of the video camera that are capable of registering the thickness of accumulated debris or sediment at the shaft excavation. The system is capable of testing shafts with depths in excess of 200 feet. Several generations of this device exist that range in size from less than a foot in diameter to over 3 feet in diameter. The inspection is viewed in realtime on a color video monitor and recorded on a standard VHS tape. Voice annotations are recorded simultaneously during the inspection process similar to standard camcorders.

Crosshole Sonic Logging (CSL)

Crosshole sonic logging is a geophysical test method used to determine the compression wave velocity between two parallel, water-filled tubes or slurry filled boreholes. By using two geophones (one emitting and one receiving) the sound wave arrival times can be logged at various depths within the tubes. From this information the insitu properties of the materials between the tubes can be inferred thus identifying various strata. More recently, this test has become a non-destructive method for evaluating the quality of newly placed drilled shaft concrete. Therein, the arrival times are measured between logging tubes attached peripherally to the reinforcing cage allowing concrete quality between the tubes to be assessed. As only the concrete in a direct line between the tubes can be tested, multiple access tubes can be installed. Typically, one tube for every foot of diameter is required to satisfactorily survey a representative portion of the shaft concrete. Data is viewed in the field on a special data acquisition system.

Shaft Integrity Test (SIT) or Pile Integrity Test (PIT)

The shaft/pile integrity test is an impact echo test that uses the reflections of anomalous cross sectional shaft or pile dimensions to determine the quality of a drilled shaft, auger-cast-in-situ or driven pile. The reflected sound waves from within the concrete are plotted as a function of arrival times which can then be correlated to the depth from which the reflection emanated. Figures SIT-1 and 2 shows the equipment used to conduct the test as well as the outputted results.

This test is well-suited for determining the depth of the foundation as well as the depth to anomalous features. However, it can not determine the magnitude of anomalous features, it requires access to the pile top to minimize confounding signals, and it is generally limited to depths on the order of 50 times the pile diameter.

Pile Driving Analyzer (PDA)

During pile driving, the stresses and accelerations imparted to the pile can be monitored and recorded to assess the quality of the installation. Although this information is also used to ascertain the load carrying capability of the pile, the quality assurance associated with type of equipment is perhaps its greatest contribution. Therein, the tensile and compressive stresses in piles can be monitored via strain gage instrumentation to prevent unnecessary damage while adjusting pile driving hammer energy to maximize production rates. The movement is also monitored using integrated accelerometer data. Figure 22 shows the instrumentation and its position during pile driving.

Post Grout Test

The post grout test is a bi-product of an end bearing enhancement technique used during the construction of drilled shafts. This test is relatively simple in concept yet confirms the performance of every grouted shaft up to a lower limit of shaft capacity. During the process of tip grouting, the upward displacement, grout pressure, and grout volume are recorded. This information provides the design engineer the response of the shaft to loading. Therein, the side shear as well as the end bearing of the shaft are verified up to the level of the applied grout pressure. The product of the grout pressure and tip area produce the tip load; this preloading is afforded by an equivalent reaction from the side shear component. Therefore, the proven capacity of the shaft is established as twice the tip load. The upper limit of capacity can be shown to be on the order of 2 to 3 times the proven capacity when verified by downward load testing. The design of post grouted shafts is discussed in a later chapter. Figure PG1-2 show the standard field data obtained from every grouted shaft. Figure PG3 shows the performance for each of 76 shafts grouted on a bridge project in West Palm Beach, Florida.

Load Testing

Load testing is considered to be the best measure of foundation suitability to resist anticipated design loads. These tests involve the application of a load capable of displacing the foundation and determining its capacity from its response. Various approaches have been devised to obtain this information. When comparing these approaches, they can be sorted from simplest to most complex in the following order: static load test, rapid load test, and the dynamic load test. These categories can be delineated by comparing the duration of the loading event with respect to the axial natural period of the foundation (2L/C), where L represents the foundation length and C represents the strain wave velocity. Test durations longer than 1000 L/C are considered static loadings and those shorter than 10 L/C are consider dynamic (Janes et al., 2000; Kusakabe et.al, 2000). Tests with a duration between 10L/C and 1000 L/C are denoted as rapid load tests. The static and rapid load test will be discussed in this chapter. The dynamic load test is discussed in the pile driving and analysis chapter.

Static Load Test

Static load testing encompasses all test methods that systematically apply an increasing load to a foundation in multiple loading increments at such a rate so as to produce no dynamic movements as stated above. These tests include many applications (i.e. deep foundations or shallow foundations, tension or compression loads) with numerous loading configurations. With regard to full-scale insitu load tests, several test procedures are most prominent: plate load test (ASTM D 1195), pile load test in compression (ASTM D 1143), pile load test in tension (ASTM D 3689), and the Osterberg load test.

Kentledge Load Test. The load from structures to the foundations can be compression (downward), tension (upward), or lateral (sideways). The downward load carrying resistance of a foundation encompasses most of the load conditions considered. In order to replicate these often enormous loads, several methodologies have been devised. The simplest form of a load test is the dead load or Kentledge method. This requires that the full test load be supplied in the form of dead weight stacked above the foundation on some framework. The framework must be capable of supporting the entire load at a single location where a hydraulic ram / jack can progressively transfer the load to the top of the foundation (Figure 26). This type of test accurately predicts the foundation response at full Kentledge load, but over estimates the stiffness of the foundation at lower loads due to the presence of the dead load overburden pressure applied to the ground surface. The practical upper limit of these tests is approximately 400 tons although physical site constraints may extend or restrict this limit drastically. Further, these tests are the most expensive and time consuming to perform from the standpoint of setup requirements. As with all static load tests, these tests are typically run in compliance with ASTM D-1143 or similar standard.

Anchored Load Tests. Static load tests with anchored reaction systems are the most common of the static load tests. These tests supply the full load to the foundation via a series of tension anchors (or adjacent deep foundations) in conjunction with a beam or truss (Figure 27). The beam must resist the load applied to the foundation by transferring it the reaction anchors which are preferably no closer than 5 diameters of the foundation (center to center spacing). The reaction anchors must not displace significantly while developing the required load. Excessive upward movement from these anchors can alter the stress field surrounding the foundation being tested and decrease the resultant ultimate capacity. Due to the constraints in designing such a reaction system, rarely does an anchored static load test exceed 1500 tons. However, anchored tests as large as 3500 tons are commonplace in some parts of the world.

The analysis of static load testing requires no more than plotting the load versus displacement response. As every foundation application can have a unique failure criteria, the design engineer must decide at what displacement the foundation capacity should be determined. In some instances this is based on a given fraction of ultimate load. In other cases, it may be based upon the some displacement offset method such as the Davisson Method or the FHWA method. With LRFD-type approaches, a fraction of the ultimate capacity is compared to the factored design load in a strength limit state, and displacement is considered separately in a service limit state. Figure 28 shows typical static load test results and three common approaches to determining capacity (a) a maximum permissible displacement usually set by structural sensitivity, (b) displacement offset method, and (c) the load at which additional displacement is obtained without increase in load.

Bidirectional Static Load Test. The Osterberg Cell (O-cell®) Test provides a simple, efficient and economical method of performing a static load test on a deep foundation. The O-cell is a sacrificial jack which the Engineer installs at the bottom of a pile or drilled shaft. It provides a static load and requires no overhead load frame or other external reaction system (Figure 29). The O-cell is easily installed in drilled shafts using common construction equipment and is attached to the tip of a driven pile before driving.

Installation methods on a drilled shaft can vary, but the following procedures are typical. An O-
cell or O-cells are attached to a top and bottom steel plate, which is then placed near or at the bottom of a shaft as part of the reinforcing cage or carrying frame (Figure 30). Strain gage instruments are also attached to the assembly and all wires are channeled to the surface via the cage or beam. The complete assembly is then lifted and set into the open shaft prior to the concreting process. In the case of multi-level O-cell assemblies, or placement of the O-cell off the bottom of the shaft, a tremie pipe is fed through a prefabricated hole in the steel plates to ensure proper cementation below the O-cells.

Once the concrete has reached required strength the O-cell is pressurized. The O-cell uses the soil system for reaction, eliminating the need for overhead or external reaction systems. The O-cell is expanded until the expansion force is some desired proof multiple of the design loading, or one of the two components, side friction or end bearing, reaches some defined failure condition, or the cell reaches its maximum expansion.

Depending on the shaft diameter, O-cells can be grouped together on a single plane to increase the effective load. Testing is typically performed following the ASTM Quick Test Method D1143 (ASTM, 1993). Instrumentation used to measure load and deflection is similar to instrumentation used for conventional load tests. At the completion of the test the cell can be filled with grout to reestablish its integrity and permit the test shaft or pile to become a production shaft or pile.

The O-cell loads the test pile in compression similar to a conventional static load test. Data from an O-cell test is therefore analyzed much the same way as conventional static test data. The only significant difference is that the O-cell provides two load-versus-movement curves, one for shaft resistance and one for toe resistance (Figure 31). The failure load for each component may be determined from these curves using a failure criteria similar to that recommended for conventional load tests. To determine the shaft resistance capacity, the buoyant weight of the pile should be subtracted from the upward O-cell load. Analysis for the toe resistance should not include elastic deformation since the load is applied directly.

The engineer may further utilize the component curves to construct an equivalent pile head load-
deflection curve and investigate the overall pile capacity. If the pile is then assumed rigid, the pile head and toe move together and have the same deflection at this load. By adding the shaft resistance to the mobilized toe resistance at the chosen deflection, a single point on the equivalent pile head load curve is determined. Additional points may then be calculated to develop the curve up to the maximum deflection (or maximum extrapolated deflection) of the component that did not fail. Points beyond the maximum deflection of the non-failing component may also be obtained by conservatively assuming that at greater deflections if remains constant at the maximum applied load.

An O-cell test can be performed on types of shafts and piles, including; pre-cast, open-ended pipe piles, mandrel driven piles, and has been used on large diameter drilled shafts. O-cells can be used to test deep foundations over water or confined areas because the O-cell test does not require an overhead reaction system. An O-cell test may be applied in many situations due to the systems flexibility, e.g. placement of the O-cell within the shaft may be altered, or additional layers of O-
cells may be used to isolate significant soil zones.

Rapid Load Test

Since its inception in1988, the inertia loading technology called Statnamic testing has gained popularity with many designers largely due to its time efficiency, cost effectiveness, data quality, and flexibility in testing existing foundations. Where large capacity static tests may take up to a week to set up and conduct, the largest of Statnamic tests (3500 tons) typically takes no more than a few days. Further, multiple smaller-capacity tests (up to 2000 tons) can easily be completed within a day. The direct benefit of this time efficiency is the cost savings to the client and the ability to conduct more tests within a given budget. Additionally, this test method has boosted quality assurance by giving the contractor the ability to test foundations thought to have been compromised by construction difficulties without significantly affecting production and without requiring previous planning for its testing.

Statnamic testing is designated as a rapid load test that uses the inertia of a relatively small reaction mass instead of a reaction structure to produce large forces. The duration of the Statnamic test is typically 100 to 120 milliseconds, but is dependant on the ratio of the applied force to the weight of the reaction mass. Longer duration tests of up to 500 milliseconds are possible but require more reaction mass.

The Statnamic force is produced by quickly-formed high pressure gases that in turn launch a reaction mass upward at up to twenty times the acceleration of gravity. The equal and opposite force exerted on the foundation is simply the product of the mass and acceleration of the reaction mass. It should be noted that the acceleration of the reaction mass is not significant in the analysis of the foundation; it is simply a by-product of the test. Secondly, the load produced is not an impact in that the mass is in contact prior to the test. Further, the test is over long before the masses reach the top of their flight. The parameters of interest are only those associated with the movement of the foundation (i.e. force, displacement, and acceleration). Figure 32 shows the setup for both an axial compression and lateral statnamic test setup.

Analysis Procedures. Typical analysis of Statnamic data relies on measured values of force, displacement and acceleration. A soil model is not required, hence, the results are not highly user dependent. The Statnamic forcing event induces foundation motion in a relatively short period of time and hence acceleration and velocities will be present. The accelerations are typically small(1-
2 g's), however the enormous mass of the foundation when accelerated resists movement due to inertia and as such the fundamental equation of motion applies, Equation 1.

where,	F = forcing event

m = mass of the foundation

a = acceleration of the displacing body

v = velocity of the displacing body

c = viscous damping coefficient

k = spring constant of the displacing system

x = displacement of the body

The equation of motion is generally described using four terms: forcing, inertial, viscous damping, and stiffness. The forcing term (F) denotes the load application which varies with time and is equated to the sum of remaining three terms. The inertial term (ma) is the force which is generated from the tendency of a body to resist motion, or to keep moving once it is set in motion (Young, 1992). The viscous damping term (cv) is best described as the velocity dependant resistance to movement. The final term (kx), represents the classic system stiffness, which is the static soil resistance.

When this equation is applied to a pile/soil system the terms can be redefined to more accurately describe the system. This is done by including both measured and calculated terms. The revised equation is displayed below:

where, FStatnamic is the measured Statnamic force, m is the calculated mass of the foundation, a is the measured acceleration of the foundation, c is the viscous damping coefficient , v is the calculated velocity, and Fstatic is the derived pile/soil static response.

There are two unknowns in the revised equation Fstatic and c, thus the equation is under specified. Fstatic is the desired value, so the variable c must be obtained to solve the equation. Middendorp (1992) presented a method to calculate the damping coefficient referred to as the Unloading Point Method (UP). With the value of c known, the static force can be calculated. This force, termed "Derived Static," represents an equivalent soil response similar to that produced by a traditional static load test.

Unloading Point Method. The UP is a simple method by which the damping coefficient can be determined from the measured Statnamic data. It uses a simple single degree of freedom model to represent the foundation/soil system as a rigid body supported by a non-linear spring and a linear dashpot in parallel (see Figure 32). The spring represents the static soil response (FStatic) which includes the elastic response of the foundation as well as the foundation/soil interface and surrounding soil response. The dashpot is used to represent the dynamic resistance which depends on the rate of pile penetration (Nishimura, 1995).

The UP makes two primary assumptions in its determination of "c." The first is the static capacity of the pile is constant when it plunges as a rigid body. The second is that the damping coefficient is constant throughout the test. By doing so a time window is defined in which to calculate the damping coefficient as shown in Figure 33. This figure shows a typical Statnamic load-
displacement curve which denotes points (1) and (2).

The first point of interest (1) is that of maximum Statnamic Force. At this point the static resistance is assumed to have become steady state, for the purpose of calculating "c". Thus, any extra resistance is attributed to that of the dynamic forces (ma and cv). The next point of interest (2) is that of zero velocity which has been termed the "Unloading Point." At this point the foundation is no longer moving and the resistance due to damping is zero. The static resistance, used to calculate "c" from point (1) to (2), can then be calculated by the following equation:

Where, F Statnamic, m, and a are all known parameters; F Static UP is the static force calculated at (2) and assumed constant from (1) to (2).

Next, the damping coefficient can be calculated throughout this range, from maximum force (1) to zero velocity (2). The following equation is used to calculate c:

Damping values over this range should be fairly constant. Often the average value is taken as the damping constant, but if a constant value occurs over a long period of time it should be used (see Figure 34).

Note that as v approaches zero at point (2), values of c can be different from that of the most representative value and therefore the entire trend should be reviewed. Finally the derived static response can be calculated as follows:

Currently software is available to the public that can be used in conjunction with Statnamic test data to calculate the derived static pile capacity using the UP Method (Garbin, 1999). This software was developed by the University of South Florida and the Federal Highway Administration and can be downloaded from www.eng.usf.edu/~gmullins under the Statnamic Analysis Workbook (SAW™) heading.

The UP has proven to be a valuable tool in predicting damping values when the foundation acts as a rigid body. However, as the pile length increases an appreciable delay can be introduced between the movement of the pile top and toe, hence negating the rigid body assumption. This occurrence also becomes prevalent when an end bearing condition exists; in this case the lower portion of the foundation is prevented from moving jointly with the top of the foundation.

Middendorp (1995) defines the "Wave Number" (Nw) to quantify the applicability of the UP. The wave number is calculated by dividing the wave length (D) by the foundation depth (L). D is obtained by multiplying the wave speed c in length per second by the load duration (T) in seconds. Thus, the wave number is calculated by the following equation:

Through empirical studies Middendorp determined that the UP would predict accurately the static capacity from Statnamic data, if the wave number was greater than 12. Nishimura (1995) established a similar threshold at a wave number of 10. Using wave speeds of 5000 m/s and 4000 m/s for steel and concrete respectively and a typical Statnamic load duration, the UP is limited to piles shorter than 50 m (steel) and 40 m (concrete). Wave number analysis can be used to determine if stress waves will develop in the pile. However, this does not necessarily satisfy the rigid body requirement of the UP.

Statnamic tests cannot always produce wave numbers greater than 10, and as such there have been several methods suggested to accommodate stress wave phenomena in Statnamically tested long piles (Middendorp, 1995). Due to limitations on paper length these methods are not presented.

Modified Unloading Point Method. Given the limitations of the UP, users of Statnamic testing have developed a remedy for the problematic condition that arises most commonly. The scenario involves relatively short piles (Nw>10) that do not exhibit rigid body motion, but rather elastically shorten within the same magnitude as the permanent set. This is typical of rock-socketed drilled shafts or piles driven to dense bearing strata that are not fully mobilized during testing. The consequence is that the top of pile response (i.e. acceleration, velocity, and displacement) is significantly different from that of the toe. The most drastic subset of these test results show zero movement at the toe while the top of pile elastically displaces in excess of the surficial yield limit (e.g. upwards of 25 mm). Whereas with plunging piles (rigid body motion) the difference in movement (top to toe) is minimal and the average acceleration is essentially the same as the top of pile acceleration; tip restrained piles will exhibit an inertial term that is twice as large when using top of pile movement measurements to represent the entire pile.

The Modified Unloading Point Method (MUP), developed by Justason (1997), makes use of an additional toe accelerometer that measures the toe response. The entire pile is still assumed to be a single mass, m, but the acceleration of the mass is now defined by the average of the top and toe movements. A standard UP is then conducted using the applied top of pile Statnamic force and the average accelerations and velocities. The derived static force is then plotted versus the top of pile displacement as before. This simple extension of the UP has successfully overcome most problematic data sets. Plunging piles instrumented with both top and toe accelerometers have shown little analytical difference between the UP and the MUP. However, MUP analyses are now recommended whenever both top and toe information is available.

Although the MUP provided a more refined approach to some of the problems associated with UP conditions, there still exists a scenario where it is difficult to interpret Statnamic data with present methods. This is when the wave number is less than ten (relatively long piles). In these cases the pile may still only experience compression (no tension waves) but the delay between top and toe movements causes a phase lag. Hence an average of top and toe movements does not adequately represent the pile.

SUP Method. The fundamental concept of the Segmental Unloading Point (SUP) method is that the acceleration, velocity, displacement, and force from each segment of a pile can be determined using strain gage measurements along the length of the pile (Mullins, et al., 2002). Individual pile segment displacements are determined using the relative displacement as calculated from strain gage measurements and an upper or lower measured displacement. The velocity and acceleration of each segment are then determined by numerically differentiating displacement then velocity with respect to time. The segmental forces are determined by calculating the difference in force from two strain gage levels.

Typically the maximum number of segments is dependent on the available number of strain gage layers. However, strain gage placement does not necessitate assignment of segmental boundaries; as long as the wave number of a given segment is greater than 10, the segment can include several strain gage levels within its boundaries. The number and the elevation of strain gage levels are usually determined based on soil stratification; as such, it can be useful to conduct an individual segmental analysis to produce the shear strength parameters for each soil strata. A reasonable upper limit on the number of segments should be adopted because of the large number of mathematical computations required to complete each analysis. Figure 36 is a sketch of the SUP pile discretization.

The notation used for the general SUP case defines the pile as having m levels of strain gages and m+1 segments. Strain gage locations are labeled using positive integers starting from 1 and continuing through m. The first gage level below the top of the foundation is denoted as GL1 where the superscript defines the gage level. Although there are no strain gages at the top of foundation,this elevation is denoted as GL0. Segments are numbered using positive integers from 1 to m+1, where segment 1 is bounded by the top of foundation (GL0) and GL1. Any general segment is denoted as segment n and lies between GLn-1 and GLn. Finally, the bottom segment is denoted as segment m+1 and lies between GLm and the foundation toe.

Calculation Of Segmental Motion Parameters. The SUP analysis defines average acceleration, velocity, and displacement traces that are specific to each segment. In doing so, strain measurements from the top and bottom of each segment and a boundary displacement are required. Boundary displacement may come from the Statnamic laser reference system (top), top of pile acceleration data, or from embedded toe accelerometer data.

The displacement is calculated at each gage level using the change in recorded strain with respect to an initial time zero using Equation 7. Because a linearly-varying strain distribution is assumed between gage levels, the average strain is used to calculate the elastic shortening in each segment.

Level displacements

where

x n	= the displacement at the nth gage level

average seg n	= the average change in strain in segment n

L seg n	= the length of the nth segment

To perform an unloading point analysis, only the top-of-segment motion needs to be defined. However, the MUP analysis, which is now recommended, requires both top and bottom parameters. The SUP lends itself naturally to providing this information. Therefore, the average segment movement is used rather than the top-of-segment; hence, the SUP actually performs multiple MUP analyses rather than standard UP. The segmental displacement is then determined using the average of the gage level displacements from each end of the segment as shown in the following equation:

where x seg n is the average displacement consistent with that of the segment centroid.

The velocity and acceleration, as required for MUP, are then determined from the average displacement trace through numerical differentiation using Equations 9 and 10, respectively:

where	vn	=	the velocity of segment n

an	=	the acceleration of segment n

t	=	the time step from time t to t+1

It should be noted that all measured values of laser displacement, strain, and force are time dependent parameters that are field recorded using high speed data acquisition computers. Hence the time step, t, used to calculate velocity and acceleration is a uniform value that can be as small as 0.0002 seconds. Therefore, some consideration should be given when selecting the time step to be used for numerical differentiation.

The average motion parameters (x, v, and a) for segment m+1 can not be ascertained from measured data, but the displacement at GLm can be differentiated directly providing the velocity and acceleration. Therefore, the toe segment is evaluated using the standard UP. These segments typically are extremely short (1 - 2 m) producing little to no differential movement along its length.

Segmental Statnamic and Derived Static Forces. Each segment in the shaft is subjected to a forcing event which causes movement and reaction forces. This segmental force is calculated by subtracting the force at the top of the segment from the force at the bottom. The difference is due to side friction, inertia, and damping for all segments except the bottom segment. This segment has only one forcing function from GLm and the side friction is coupled with the tip bearing component. The force on segment n is defined as:

where	Sn	=	the applied segment force from strain measurements

En	=	the composite elastic modulus at level n

An	=	the cross sectional area at level n

n	=	the measured strain at level n

Once the motion and forces are defined along the length of the pile, an unloading point analysis on each segment is conducted. The segment force defined above is now used in place of the Statnamic force in Equation 2. Equation 12 redefines the fundamental equation of motion for a segment analysis:

where,	S n Static	= the derive static response of segment n

m n	= the calculated mass of segment n

c n	= the damping constant of segment n

The damping constant (in Equation 13) and the derived static response (Equation 14) of the segment are computed consistent with standard UP analyses:

Finally the top-of-foundation derived static response can be calculated by summing the derived static response of the individual segments as displayed in the following equation:

Software capable of performing SUP analyses (SUPERSAW™) has been developed at the University of South Florida in cooperation with the Federal Highway Administration (Winters, 2002). It can be downloaded from www.eng.usf.edu/~gmullins under the Statnamic Analysis Software heading.

General References

ASTM, 1998, "Soil and Rock," Annual Book of ASTM Standards, Volume 4.08, West Conshohocken, PA.

Bloomquist, David and Townsend, Frank C., 1991, Development of Insitu Equipment for Capacity Determinations of Deep Foundations in Florida Limestone," Florida Department of Transportation Funded Research, Project #23385-001, March.

Bermingham, P., and White, J., (1995), "Pyrotechnics and the Accurate of Prediction of Statnamic Peak Loading and Fuel Charge Size", First International Statnamic Seminar, 1995, Vancouver, British Columbia Canada

Dapp, S., and Mullins, G., (2002). "Pressure-Grouting Drilled Shaft Tips: Full-Scale Research Investigation for Silty and Shelly Sands," Deep Foundations 2002: An International Perspective on Theory, Design, Construction, and Performance, ASCE Geo Institute, GSP No.116, Vol. I, pp. 335-350.

FDOT, 1999, Soils and Foundations Manual, State Materials Office, Gainesville, FL.

FHWA, 1993, Soils and Foundations, Workshop Manual - Second Edition, FHWA HI-88-009, National Highway Institute, NHI Course No. 13212, Revised July.

Garbin, E. J., (1999), "Data Interpretation for Axial Statnamic Testing and the Development of the Statnamic Analysis Workbook," Master's Thesis, University of South Florida, Tampa, FL.

Goodman, Richard E., 1989, "Rock Mechanics," 2nd ed., John Wiley and Sons, New York.

Janes, M.C., Justason, M.D., Brown, D.A., (2000), "Long period dynamic load testing ASTM standard draft," Proceedings of the Second International Statnamic Seminar, Tokyo, October, 1998, pp. 199-218.

Justason, M.D., (1997), "Report of Load Testing at the Taipei Municipal Incinerator Expansion Project," Taipei City, Taipei.

Kulhawy, F. H. and Mayne, P. W., (1990), " manual on Estimating Soil Properties for Foundation Design," EPRI EL - 6800 Research Project 1493-6, Electric Power Research Institute, August.

Kusakabe, Kuwabara, and Matsumoto (eds), (2000), "Statnamic Load Test," Draft of 'method for rapid load test of single piles (JGS 1815-2000),' Proceedings of the Second International Statnamic Seminar, Tokyo, October 1998 pp. 237-242.

Lai, P.W., 1999, "Determination of Design Skin Friction for Drilled Shafts Socketed in the Florida Limestone," Soils and Foundations Manual, State Materials Office, Gainesville, FL.

Lewis, C.L., (1999), "Analysis of Axial Statnamic Testing by the Segmental Unloading Point Method," Master's Thesis, University of South Florida, Tampa, FL.

McVay, M.C., Townsend, F.C., and Williams, R.C., 1992, "Design of Socketed Drilled Shafts in Limestone," Journal of Geotechnical Engineering, Vol. 118, No 10, October.

Middendorp, P. and Bielefeld, M.W., (1995), "Statnamic Load Testing and the Influence of Stress Wave Phenomena," Proceedings of the First International Statnamic Seminar, Vancouver, Canada, pp. 207-220.

Middendorp, P., Bermingham, P., and Kuiper, B. , (1992), "Statnamic Load Testing Of Foundation Pile." Proceedings, 4th International Conference On Application Of Stress-Wave Theory To Piles, The Hague, pp. 581-588.

Mullins, G., Dapp, S., and Lai, P., (2000). "Pressure Grouting Drilled Shaft Tips in Sand". New Technological and Design Developments in Deep Foundations, Dennis, N. D., et al. (ed.), ASCE, Geo Institute, Vol. 100, pp 1-17.

Mullins, G., Lewis, C., and Justason, M., (2002). "Advancements in Statnamic Data Regression Techniques," Deep Foundations 2002: An International Perspective on Theory, Design, Construction, and Performance, ASCE Geo Institute, GSP No.116, Vol II, pp. 915-930.

Mullins, G. and O'Neill, M. W., (2003). "Pressure Grouted Drilled Shafts," Report submitted to A.H. Beck Foundation, Inc., June, 299 pgs.

Nishimura, S., and Matsumoto, T., (1995), "Wave Propagation Analysis During Statnamic Loading of a Steel Pipe Pile," Proceedings of the First International Statnamic Seminar, Vancouver, Canada, September.

Osterberg, J.O. (1994) Recent Advances in Load Testing Driven Piles and Drilled Shafts using the Osterberg Load Cell Method. Geotechnical Lecture Series, Geotechnical Division of the Illinois Section, ASCE, Chicago IL.

Winters, D. (2002), "SUPERSAW Statnamic Analysis Software," Master's Thesis, University of South Florida, Tampa, FL, May.

Wittke, Walter, 1990, "Rock Mechanics, Theory and Application with Case Histories," Springer-
Verlag, Berlin.

Young, Hugh. D., (1992), "University Physics," Eight Edition, Addison Wesley.